By Gregor Wollmann, PH.D., P.E., M.ASCE, Lisa Brigg S, S.E., M.ASCE, Savas Kiriakidis, P.E., M.ASCE, AND Chris Daigle, P.E.
The replacement of the John Greenleaf Whittier Memorial Bridge was one of the signature projects of the Massachusetts Department of Transportation's $3-billion accelerated bridge construction program. The network tied arch structural system selected for the main span of the new northbound and southbound structures that the bridge comprises was a novel solution developed to overcome significant construction challenges.
Named for the prominent poet and abolitionist who was once a resident of Amesbury, Massachusetts, the existing John Greenleaf Whittier Memorial Bridge, more commonly referred to as the Whittier Bridge, was a truss and through-arch structure. Connecting Newburyport, Massachusetts, and Amesbury, it was opened in 1951 and carried Interstate 95 across the Merrimack River.
By the early 2000s, however, the bridge had various structural and functional deficiencies; its replacement was central to the improvement of a 4 mi section of I-95 to the north and south of the bridge. The project, led by the Massachusetts Department of Transportation, was awarded in February 2013 as a design/build contract to the joint venture of Walsh Construction and McCourt Construction, headquartered in Chicago and Boston, respectively. The Boston and New York City offices of HNTB Corp. served as the engineer of record, and Genesis Structures Inc., which has its headquarters in Kansas City, Missouri, was the erection engineer.
Two parallel network tied arch bridges-one for northbound traffic and the other for southbound traffic-were chosen to replace the existing bridge. This structure type was selected because it echoes the distinctive shape of the original bridge and because it has the structural efficiency to span the required 480 ft across the main navigation channel. The tied arches have circular ribs that rise to 74.8 ft above the tie level, resulting in a span-to-rise ratio of 6.4.
The northbound bridge was opened in November 2015 in a temporary configuration that carried three lanes of traffic in each direction. The existing bridge was demolished to make room for the southbound bridge, which was completed in December 2017. In the final configuration, each bridge accommodates four travel lanes, a 12 ft wide shoulder, and a breakdown lane. In addition, the northbound bridge features a shared-use path for pedestrian and bicycle traffic.
A network tied arch is characterized by its system of inclined hangers in which some of the hangers cross over at least two other hangers. Credit for the development and promotion of the modern network tied arch belongs to Per Tveit, a Norwegian engineer and researcher who developed this structural form in the 1950s and has been investigating it since. Despite its advantages, it is only within the past decade that network tied arches have found widespread acceptance and application in the United States.
To illustrate the effectiveness of the simple modification from a vertical to an inclined hanger arrangement, figures 1-4 on page 66 show a comparison of deflections and bending moments for these two configurations for the northbound structure. Other than the hanger inclination, the two systems are identical. Arches are very effective in carrying uniform loads, but they are sensitive to partial loading conditions. Therefore, the comparisons consider the scenario in which there are three lanes of live load over half the span length. With the network hanger arrangement, the maximum live load deflection is reduced by a factor of 5 (2.75 in. versus 13.5 in.), and the maximum tie-girder bending moment is reduced by a factor of nearly 4 (3,883 kip-ft versus 15,108 kip-ft).
Additionally, the network tied arch system is much more resilient if a tie girder is damaged. This is demonstrated in figures 3 and 4, which compare deflections and bending moments with a flexural hinge introduced into the tie girder directly under the truck load, representing loss of moment capacity because of a local plate fracture. With the network hanger arrangement, the increase in deflection from 2.75 in. to 3.70 in. is minor. The maximum bending moment in the tie girder is reduced from 3,883 kip-ft to 1,314 kip-ft, and the bending moment in the rib remains small (614 kip-ft).
This favorable response to the loss of tie-girder flexural capacity can be explained by the truss-like behavior of the network tied arch. Similar to continuous truss chords, the bending moments in an arch rib and tie girder are induced primarily by the compatibility of vertical deflections. The structure remains in equilibrium even without the ability to resist these moments. Introduction of the hinge into the tie girder softens the flexural bending load path, and the structure shifts to the stiffer truss response.
In contrast, loss of tie-girder moment capacity with the vertical hanger arrangement has an adverse impact on the structure. The resulting maximum deflection of 34.7 in. is nearly 10 times greater than with the network hanger arrangement (figure 3). Also, a portion of the bending moment released from the tie girder is shifted into the arch rib, leading to a more than 15 times larger arch-rib bending moment than with the network hanger configuration (9,400 kip-ft vs. 614 kip-ft) (figure 4). The reason for this behavior is that flexural capacity of an arch rib or a tie girder is required to satisfy equilibrium, as a truss mechanism is not available with the vertical hanger system.
Beyond enabling truss-like behavior, close hanger spacing and multiple crossovers are additional benefits of inclined hangers. Such an arrangement makes network tied arches extraordinarily resilient to the loss of one or several hangers, even when dynamic amplification because of a sudden removal is accounted for. For analysis of the sudden hanger loss load case, the Post-Tensioning Institute (PTI) publication Recommendations for Stay Cable Design, Testing, and Installation recommends an equivalent static impact factor of 100 percent of the force in the removed cable. However, it is beneficial to perform a true dynamic analysis, as this results in much lower dynamic amplification, typically ranging from 45 to 80 percent. Figure 5 shows a snapshot from such a time-dependent analysis for the northbound structure of the Whittier Bridge.
For highway bridges with spans of up to 500 ft, an open H section is feasible for the arch ribs. For longer spans, railway loading, or systems without upper lateral bracing, typically a box or other type of closed section is necessary. The advantages of an open section are simplicity in fabrication, in-field splicing of rib segments, and the convenient hanger connections via crossbeams or "fin plates" that span between the flanges (figure 6 on page 68). However, the open section has very little torsional stiffness. Therefore, it was important for the connection between the upper lateral bracing and rib to provide sufficient restraint to prevent torsional rib buckling. For the twin structures, this was accomplished by adding a pair of diaphragms (A and B in figure 7 on page 68), thus creating a connection that transfers both moment and shear into the bracing members while also stiffening the arch rib locally.
The decks are supported by 34 hangers per side. The hangers are composed of 10 or 11 greased and sheathed 0.62 in. diameter strands, grade 270 ksi. The strands are anchored via wedges in the anchor plates at each end of the cable. The system was designed to provide full encapsulation from anchor to anchor and passed stringent laboratory fatigue and water tightness tests per the PTI's recommendations.
At the upper terminal of each cable, the anchor plate bears on a cast steel fork socket, which is connected by a pin to the fin plate. While this configuration results in very simple connection details, it was important to take into account incidental forces acting perpendicular to the fin plate. Such forces may be because of a minor misalignment of the hanger, but most significant are fatigue-critical effects from wind- and traffic-induced vibrations. Because of the complex load path and stress distribution, a detailed, finite element model was developed to analyze the connection. The fin plate was made 2.5 in. thick with strong fillet welds (0.5 in. legs) to the end plates to keep fatigue stresses acceptably low.
Each tie girder is composed of four individual plates that are bolted together along their edges using rolled-angle shapes for the connection. Plate sizes were selected so that even with the complete loss of any one plate, the remaining tie section could resist all axial forces and bending moments in effect prior to the plate fracture. The eccentricity of the axial force with respect to the shifted neutral axis of the locally effective reduced section causes a large moment increase that was accounted for in the design stage. Although current standard practice, this approach is likely to be quite conservative, considering the ability of the network tied arch system to absorb the released tie-girder bending moment into the much more effective truss behavior.
The internal tie girder dimensions are 4 ft by 6 ft 2 in. The width of 4 ft has been found to be optimal for box-shaped tie girders, and it is a practical minimum needed to accommodate internal diaphragms that have sufficiently sized access openings. Increasing the width of the tie girder typically offers no advantages because it drives up the corresponding arch-rib dimension in the knuckle joint region, where plate slenderness effects prevent a commensurate increase in strength. The hangers were guided through openings in the upper flange of the tie girder and are anchored in a crossbeam spanning between its webs. The multistrand hanger system allowed for stressing one strand at a time using compact and lightweight stressing equipment.
The knuckle joint is the region at which the tie girder, arch rib, and end floor beam meet as well as where the bearing force is introduced into the structure. The end floor beam provides moment restraint to the arch ribs, thus reducing their buckling lengths in the critical portal frame region. Bearing replacement presents a challenging consideration that usually controls the design of the connection. A system of diaphragms in figure 8 on page 68 was added within the knuckle joints to accomplish the load transfer through the joint. Diaphragms A and C establish moment and shear continuity with the box-shaped end floor beam, and they assist diaphragm B as bearing stiffeners. Diaphragm D transfers the vertical component of the arch rib's axial force to the webs of the tie girder and the rib web's shear force to the flanges of the tie girder.
Each bridge deck uses 9.5 in. thick precast, posttensioned deck panels that span 12 ft between floor beams. The deck slab is composite with the floor beams, edge girders, and two light longitudinal stringers, which aided constructability. The decks are protected by a spray-on waterproofing membrane and a 3 in. thick Superpave asphaltic wearing surface.
Because of the composite action, the deck slab of each bridge participates in resisting tension in the tie girder. To counteract that tension, the deck panels are posttensioned in the longitudinal direction, which provides a nominal average prestress of 430 psi. However, much of that precompression was lost initially because of force bleeding into the tie girder; long term it will be lost because of creep and shrinkage of the concrete. Therefore, generous mild reinforcement amounting to 1.5 percent of the deck slab cross section was provided in the longitudinal direction. The combination of the prestressing, mild reinforcement, and waterproofing systems resulted in a very durable deck.
A concern with tied arches is the fracture-critical nature of the tie girders. A component is defined as fracture critical if it is in tension and its failure is expected to result in collapse of the bridge or the inability of the bridge to perform its function. For the Whittier Bridge, multiple levels of redundancy protect the tie girders; the stitched cross section provides internal redundancy. The truss-like response of the inclined hanger arrangement allows the tie girder to tolerate significant damage. Considering only the axial force, the tie-girder cross section alone provides a capacity of more than twice the demand from factored loads. Including the resistance of rebar and posttensioning tendons in the deck and the stitch angles in the tie girder, the capacity-to-demand ratio increases to 4.
There were numerous challenges during construction, including:
- tide cycle variations of +/- 9 ft on a normal day, leaving some areas of the river bottom exposed
- maintaining a 150 ft navigable channel at all times
- maintaining 100 percent of the existing traffic lanes (except for limited rolling lane closures)
- environmentally sensitive wetlands along each bank of the river, directly below, and extending beyond the footprint of the proposed bridges
- unstable steep slopes along the riverbanks where the abutment structures were to be located
- limited crane access from land and water
- significant crane capacity restrictions
- limited available area and access for material staging and storage
Given the unique site challenges, a launched girder system supporting a pair of overhead gantry cranes was used. This system allowed for significant flexibility during the erection process and served multiple functions during each phase of construction. The launched girders enabled access to the approach staging area, permitting the gantry cranes to unload steel members directly from the delivery trucks and transport them from the staging area to the final installation site. Once the plate girders for the northbound approach spans and the floor system for the main span were erected in this manner, the gantry cranes were used to deliver and assemble a 200-ton crane to erect the arch ribs. Upon completion of the northbound structure, the launched girder system serviced the old bridge during the demolition phase, enabling the contractor to remove the existing truss in large segments, which significantly expedited this phase of construction. Finally, the southbound structure was erected in the same manner as the northbound structure.
The launched girder system, however, proved to be a challenge as well. To minimize the number of falsework towers required on the project, the launched girders had to span up to 160 ft between supports. This meant that not only did they have to be designed to support the gantry crane reactions over these long spans, but they also had to be able to cantilever that far during the launching process before reaching the next support point. Given the associated long unbraced lengths, a rectangular fabricated box section was chosen for the launch girders. This torsionally stiff closed section minimized the susceptibility to lateral torsional buckling and provided improved resistance to external lateral loads and to the forces induced by the gantry cranes.
Because of the long cantilevers during launching operations, tip deflections of up to 4 ft were anticipated. To counter these deflections, the launched girders were equipped with tapered sections at both ends that allowed the leading end of the launched girder to deflect under the maximum cantilever condition and then to ease back onto the rollers when pushed onto the next falsework support.
The launched girders consisted of 50 ft long segments, which, once assembled, provided an 850 ft long pathway for the gantry crane system. However, this length was not sufficient to traverse the entire width of the river valley, so the launch girders were moved in two "pushes" for each phase of the project, requiring the contractor to build the bridges in halves. The splices connecting these segments were specially designed to accommodate other aspects of the launched girder system.
First, the splice at the bottom flange had to avoid interference with the rollers used at the falsework towers to support the girders while minimizing friction during the launching process. Second, the splice at the top flange needed to accommodate the overhead gantry crane rail and the gantry bogeys as they moved over the connection. These constraints made for a unique splice detail, comprised of Grade 70 splice plates that are more than 2 in. thick to compensate for the limits imposed on the plate width.
To advance the girders during the launching process, a push/pull system was installed at their tail ends. This system was sized to overcome the friction in the rollers, along with an allowance for additional friction because of possible misalignment of the rollers along the longitudinal axis of the beam, the extra force required to push the girders along an uphill grade, and the additional grade of the nose section of the girder itself. The push/pull system also had to accommodate any braking forces required to control the girders on a downhill grade.
Similar to the launched girder system, the falsework towers performed multiple functions. They were detailed so that they could be relocated and reconfigured during the three phases of the project-northbound bridge construction, existing bridge demolition, and southbound bridge construction. The falsework towers carried the launched girders and the gantry crane system during all phases of erection and demolition; they also supported the approach girders and the main-span floor system as they were being constructed. The need to simultaneously resist the weight of the launched girder system and the new bridges resulted in vertical design forces of up to 2,000 kips. Adding to the complexity of the design were water depths of as much as 30 ft to the mud line and 40 ft to rock, impact forces because of debris and ice over multiple East Coast winters, and tight space constraints to fit the towers between the new and existing structures.
The erection of the 480 ft main-span network arches took place in several stages. The tie girders and floor beam members were stick-built using the gantry crane system and supported on the falsework towers in the main navigation channel. Once the arch span deck framing was complete, a temporary crane mat system that spanned between floor beams was installed. With the crane mat in place, the gantry cranes were used to deliver and assemble a 200-ton crawler crane needed for erection of the arch ribs and the upper lateral bracing. The ribs were supported on falsework struts that transferred their weight directly to the falsework towers. The struts had jacking capability to aid in member fit-up and installation of the final arch keystone piece.
After all major arch segments were in place, the overhead gantry system was once again employed to disassemble the 200-ton crawler crane and replace it with a 100-ton telescoping boom crane that was used to install the precast deck panels and the cable hangers for the arch. The cables were stressed in a specified sequence to target forces established by the erection engineer. This sequence was selected to prevent buckling of the unbraced arch and to minimize the need for cable-length adjustments at the end of construction.
Midway through the initial cable-stressing process, the sand-jack bearings at the falsework towers were released, which freed the arch to span the full 480 ft across the navigation channel. Once the remaining cables were stressed, the deck closure placements and deck posttensioning were completed, barriers and overlays were installed, and final tuning of the cable forces was performed to conclude the erection process.
The John Greenleaf Whittier Memorial Bridge is the latest in a series of network tied arches that have been constructed in the United States over the last decade. This structural form combines the advantages of truss and arch systems and provides resiliency superior to either alone. The new bridge and adjacent work greatly improved the capacity and safety of this critical transportation corridor, and the shared-use path, opened in October, provides access to nearby existing and planned recreational trails and attractions. The innovative erection scheme was instrumental for the design/build team to win the project.
Owner Massachusetts Department of Transportation Boston
Owner's engineer WSP Boston office
Design/build contractor a joint venture of Walsh Construction, Chicago, and McCourt Construction, Boston
Engineer of record HNTB Corp., Boston and New York City offices
Erection engineer Genesis Structures Inc., Kansas City, Missouri
Steel fabricator Structal-Bridges, a Division of Canam Group Inc., Claremont, New Hampshire
Hangers Freyssinet USA Inc., Sterling, Virginia
Deck posttensioning DYWIDAGSystems International USA Inc., Bolingbrook, Illinois